Manual on Subsurface Investigations (2019)

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66 C H A P T E R 5 In Situ Testing of Soil and Rock Introduction Because the vast body of natural soil and rock at the project construction site will serve as the primary bearing medium for new bridges, highways, cut slopes, walls, and embankments, in situ geotechnical tests provide valuable information concerning the field strength, deformation properties, stress state, and hydraulic conductivity of the underlying geomaterials. The term in situ is derived from Latin meaning in its original position” and refers to the testing of geomaterials while they remain in the ground, in contrast to the more traditional approach of taking samples from boreholes and transporting them to the laboratory for testing. In situ tests are generally faster than lab testing as the results can be obtained immediately while on-site. With many in situ tests, field data are collected more abundantly than laboratory data, thus allowing for a more comprehensive definition of soil strata, zones, layering, and stratigraphy, as well as the identification of lenses, weak zones, and inclusions. Field tests also allow an investigation of vertical and horizontal variability to determine whether an overall homogeneity or heterogeneity exists across a site. In situ testing also provides an independent evaluation of geotechnical engineering parameters for analysis and design. A wide variety of in situ geotechnical tests have been developed for assessing and characterizing the ground (Broms and Flodin 1988, Robertson 1986, Mayne 2012). Figure 5-1 shows an assortment of selected devices available, along with their associated acronyms. Many of these devices are intended for a particular use and are not widely used in practice. The focus in this chapter is on in situ methods that are commonly used for obtaining undrained shear strength ( ) in clays, friction angle in sands, elastic soil modulus, preconsolidation stress, and hydraulic conductivity (K). Hybrid tests have been developed for some of the tests. Hybrid tests include more than one method in a single device, such as the CPTu that combines a penetrometer with a piezo probe. A special category of hybrid tests combines in situ geotechnical and geophysical testing. This optimizes the collection of field data economically, expediently, and efficiently. Examples of hybrid geotechnical-geophysics testing devices include the seismic piezocone penetration (SCPTu) and resistivity piezocone (RCPTu). The in situ tests presented in the following sections have been grouped into three broad categories: 1. Borehole tests in soils 2. Direct-push methods for soils 3. Rock tests

67 Source: Paul Mayne Figure 5-1. Selection of in situ tests for geotechnical site characterization Borehole Test Methods Borehole-type field tests can be conducted in open borings, beneath borehole casings, or below the bottoms of HSAs, depending upon the soil conditions and drilling methods used. 5.2.1 Standard Penetration Test One of the oldest and most widely used in situ tests is the SPT. The SPT test is generally used to estimate the shear strength parameters of sandy soils and overconsolidated clays. It is unreliable for soils containing course gravel, cobbles, boulders, cohesionless silts, and soft and sensitive clays. Figure 5-2 depicts the general layout of the SPT using a hammer system, drill rods, and split-spoon or split-barrel sampler.

68 Source: Paul Mayne Figure 5-2. SPT setup and procedures The testing procedures for the SPT are found in AASHTO T 206 (ASTM D1586). The basic testing procedure consists of driving a hollow steel tube with an outside diameter of 2.0 in. (51 mm) and an inside diameter of 1.38 in. (35 mm) into the ground a vertical distance of 18 in. (46 cm) and counting the number of blows required to drive each 6-in. (15-cm) increment. The first increment is considered a seating of the sampler. The blows obtained for the second and third increments are summed to give an N-value, which is reported in blows per foot (bpf). The test is typically conducted at 5-ft (1.5-m) depth intervals. At shallow depths of less than 10 ft (3 m), the depth intervals are often less (e.g., 2.5-ft [0.75-m] intervals). An example profile of measured SPT N-values from a boring by the Vermont Agency of Transportation (VTrans) is presented in Figure 5-3.

69 Source: Paul Mayne (based on data from VTrans) Figure 5-3. Example of SPT N-values 5.2.1.1 Correction Factors Since the introduction of the SPT in 1902, the drop hammer systems have evolved from pin weight to donut to safety to automatic hammers. Automatic hammers are now predominant in the United States. All systems have a nominal hammer weight of 140 lb. and drop height of 30 in. (76 cm), giving a theoretical energy of 4,200 pound force (lbf) per in. or 350 lbf per ft per blow if 100 percent efficiency were obtained. However, actual efficiencies can be significantly less than 100 percent and quite variable, ranging from 30 to 95 percent, depending upon hammer type, commercial system, and age and condition of the equipment. As a result, it is paramount to calibrate the energy efficiency of the SPT system (including rig, hammer, and even the crew) approximately once per year so that an energy ratio (ER) can be obtained for correcting the measured N-value in accordance with ASTM D4633. At the time that the geotechnical profession became aware that energy efficiency was important (Seed et al. 1985, Skempton 1986), the average energy efficiency of SPTs in practice was approximately 60 percent, which became the reference value for the correction. This reference value is designated as N60, and it is calculated as follows: = = 60 where CE = correction factor for hammer energy Nm = the measured blow count 0 2 4 6 8 10 12 14 16 18 20 0 20 40 60 80 100 De pt h (fe et ) SPT N-value (blows per foot) Raw (N) Corrected (N60) Auger refusal at 18.8 ft Sandy gravel Clay Gravelly sand to Sandy gravel Sandy silt

70 An example of the large differences in N-values obtained using a safety hammer and automatic hammer in two side-by-side borings advanced at the national geotechnical experimentation site at Northwestern University can be seen in Figure 5-4. Here, the SPTs are conducted in a fine sand layer that extends to a depth of approximately 22 ft (6.7 m). After correcting for hammer energy, it is evident that the N60 profiles are in good agreement for both hammer systems. However, without the correction, the N-values from the automatic hammer are too low (CE = 1.58), while those from the safety hammer are too high (CE = 0.67). (a) (b) Source: Paul Mayne Figure 5-4. Comparison of (a) uncorrected (Nm) and (b) energy-corrected (N60) SPT blow counts There is a general misconception by geotechnical engineers that SPTs conducted using an automatic hammer do not need to be corrected for hammer energy. Table 5-1 shows a compilation of thousands of hammer energy measurements (ASTM D4633) by different agencies, all using automatic hammers. The measured values of ER range from 46 to 95 percent. The results highlight the importance of correcting for hammer efficiency even for automatic hammers. The state-of-the-practice using automatic hammers has increased to a mean ER ±1 standard deviation (SD) of 82 ±7 percent based on more than 17,000 hammer energy measurements (Honeycutt et al. 2014). Table 5-1. Summary of hammer energy measurements using automatic hammers Energy Ratio (percent) Automatic Hammer System Mean Standard Deviation Location or Agency Diedrich D-120 46 ± 8.7 Utah DOT Diedrich D-50 56 ± 2 Tennessee CME 850 62.7 ± 4.0 Utah DOT BK-81 w/ AW-J rods 68.6 ± 7.5 ASCE/WA Mobile B-80 70.4 ± 4.6 Utah DOT CME hammer w/ skid 72.9 ± 4.2 Washington Diedrich D50 76 ± 5.3 Florida DOT CME 45c Skid 77.4 ± 5 VTrans Diedrich D-120 78 ± 4 Tennessee CME 55 78.4 ± 8.2 Florida DOT

71 Energy Ratio (percent) Automatic Hammer System Mean Standard Deviation Location or Agency CME 850 79 ± 2 Tennessee CME 45c Track 80.6 ± 3.9 VTrans CME 45 80.7 ± 10.1 Florida DOT CME 45c Track 81.1 ± 5.8 VTrans CME 85 81.2 ± 3.9 Florida DOT CME 75 w/ AW-J rods 81.4 ± 4.7 ASCE/WA CME 75 83.1 ± 5.1 Florida DOT CME 75 Track 84 ± 5.3 VTrans CME 55 Track 85 ± 4.9 VTrans CME 750 86.6 ± 6.2 Utah DOT CME 55 Track 87.4 ± 5.4 VTrans Mobile B-57 88 ± 3 Tennessee Mobile B-57 93 ± 3 Tennessee CME 75 rig 94.6 ± 2.1 Utah DOT Sources: ASCE/WA (Batchelor et al. 1995), Florida DOT (Davidson et al. 1999), VTrans (Kelley and Lens 2010), Utah DOT (Sjoblom et al. 2002), and Tennessee (Webster and Rawlings 2009) In addition to hammer energy (CE), the N-value is also affected by factors such as borehole diameter (CB), rod length and type (CR), and sampler configuration (CS). Therefore, the corrected N60 is calculated by adjusting the measured blow counts using the following relation: = Approximate values for the various correction factors are given in Table 5-2. Table 5-2. SPT correction factors for field procedures Factor Influencing Variable Field Case Factor Values CN Depth effect due to increasing effective overburden stress (σvo') σatm = 1 atmosphere = 1.013 bars = 101.3 kilopascals = 1.058 tsf = 14.7 psi CN = (σatm/σvo')0.5 ≤ 2 CE CE = ER/60 where ER = hammer energy ratio (ASTM D4633) Hammer Type: Automatic: Safety: Donut: Pinweight: Range of Factor CE 1.0 to 1.6 0.8 to 1.3 0.6 to 0.8 0.5 to 0.7 CB Borehole diameter, b (in.) 2.5 < b ≤ 4.5 b = 6 b = 8 CB = 1.00 CB = 1.05 CB = 1.15 CS* Split-barrel sampler With liner: No liner: CS = 1.0 CS = 1.2

72 Factor Influencing Variable Field Case Factor Values CR Drill rod length, L (ft) L > 33 20 < L< 33 13 < L < 20 10 < L < 13 L < 10 CR = 1.00 CR = 0.95 CR = 0.85 CR = 0.80 CR = 0.75 Source: modified from Skempton (1986), Kulhawy and Mayne (1990), Youd et al. (2001) * Common United States practice is no liner. psi: pounds per square inch tsf: tons per square foot 5.2.1.2 Stress Normalization For clean sands, there can be a stress normalization of N60, often called an overburden correction, which references N60 to an equivalent effective stress level of one atmosphere (1 atm), designated (N1)60 and calculated as follows: ( ) = A variety of stress normalization factors (CN) have been developed (e.g., Kulhawy and Mayne 1990, Boulanger and Idriss 2014, ASTM D6066). A simple and commonly accepted relationship is as follows: = ( ⁄ ) . ≤ 2 where = effective vertical overburden stress = 1 atm ≈ 1 bar = 100 kilopascals (kPa) = 14.7 psi ≈ 1 ton per square foot (tsf) ≈ 1 kilogram per square centimeter (kg/cm2) 5.2.2 Vane Shear Test VSTs have been used since the 1940s for evaluating undrained shear strength and sensitivity of clays and silts in situ (Chandler 1988). The test involves a four-sided blade (vane) that is pushed into the soil deposit, usually at 3.3-ft (1-m) depth intervals (Figure 5-5). Within one minute after blade insertion, a moment is applied to twist the rods at 0.1 degrees per second to cut a cylindrical section of clay (ASTM D2573). The peak torque reading ( ) is used to determine the maximum shear stresses acting on the soil cylinder, which is interpreted as the vane undrained shear strength ( ). The VST is usually applied to clays where < 4 ksf (200 kPa).

73 Source: Paul Mayne Figure 5-5. VST or field van for clays and silts The standard vane is rectangular with a blade width (D) of 65 mm (2.65 in.) and height (H) 130 mm (5.12 in.), giving a height-to-width ratio of H/D = 2, with a blade thickness of 2 mm (0.08 in.). Other vane configurations and sizes are available to optimize the range of the load cells and torque gages, depending upon ground conditions (Chandler 1988). For example, tapered blades facilitate insertion (and extraction) in stiffer soils. Figure 5-6 presents the general solution for calculating suv for vanes with any D, H, and taper angles at top ( ) and bottom ( ), as well as the most common geometries.

74 Source: Paul Mayne Figure 5-6. Limit equilibrium solutions for calculating undrained shear strength of rectangular and tapered vanes Modern field vanes employ downhole electronic torque meters and rheometers to measure moment and rotation of the blades (Peuchen and Mayne 2007). The data are obtained in either analog or digital format and then transmitted uphole using an electronic cable for data logging onto a field computer. With modern data acquisition systems, the electrovane shear test (EVST) provides continuous records of the entire torque vs. angle of rotation response for the test (not just peak and residual torque). This data can be interpreted to provide a shear stress vs. shear strain relationship. The shear stress vs. shear strain relationship can subsequently be used to estimate the stiffness of the soil. Example results of the EVST are illustrated in Figure 5-7. Source: Paul Mayne Figure 5-7. EVST results from selected depths in clay deposit If desired, an additional 10 revolutions are performed (faster rate permitted during this phase), and, subsequently, the residual torque reading is recorded at the aforementioned standard rate to evaluate the remolded undrained shear strength ( ). The ratio of peak-to-remolded strengths (at the same water content) defines the in-place sensitivity, = ⁄ .

75 Example results of EVST in soft clay for a bridge and embankment crossing in Idaho are presented in Figure 5-8. The peak values of undrained shear strength are seen on the right side of the figure, averaging around = 777 psf (37.2 kPa), while the remolded strengths are seen on the left side, averaging 182 pounds per square foot (psf) (8.7 kPa). The sensitivity of the clay has a mean value of = 4.3. Source: Paul Mayne (based on data from Idaho Transportation Department) Figure 5-8. Example of EVST in soft clay 5.2.3 Pressuremeter Testing The pressuremeter testing (PMT) is primarily used to estimate the stiffness and the shear strength of soils. The testing process involves lowering the pressuremeter to the desired test depth, inflating the cylindrical rubber balloon by injecting fluid, and measuring the associated changes in pressure and volume of the injected fluid. Tests are usually conducted at 3.3-ft (1-m) depth intervals. For soils, the injected fluid is typically water, while in hard soils and rocks, hydraulic oil is typically used. It is also feasible to use gas or air pressure while testing soils, especially with computerized PMT systems. The components of the PMT are illustrated in Figure 5-9.

76 Source: Paul Mayne Figure 5-9. Basic components and procedures of the PMT Figure 5-10 shows an example of a pressure vs. volume curve obtained from a PMT conducted in sand. The recorded pressure-volume change response allows the derivation of an entire stress-strain-strength curve and the evaluation four specific soil parameters at each test depth. The four soil parameters include (i) lift-off pressure ( ), (ii) stiffness or modulus, (iii) shear strength, and (iv) limit pressure . Source: Florida DOT Figure 5-10. Example PMT pressure-volume curve

77 The lift-off pressure is considered to relate to the geostatic horizontal total stress (σh0) and may be used to estimate the lateral stress coefficient, = ⁄ , where the effective horizontal stress = − . Depending on the data reduction method adopted, the stiffness is expressed as either a Young's modulus ( or ) or shear modulus ( or ), where the subscript "u" indicates an assumed undrained condition at constant volume, and the prime indicates a fully drained condition. The elasticity relationship =2 (1 + ) can be applied to relate the two moduli where ѵ = Poisson's ratio: = 0.5 (undrained) and ≈ 0.2 (drained). In Figure 5-10, an unload-reload cycle was performed at an applied pressure of 111 psi to better define the elastic modulus. The shear strength is evaluated as either the undrained shear strength ( ) for undrained loading (i.e., constant volume), or the effective friction angle ( ) for fully drained conditions (i.e., no excess pore pressure). Finally, the limit pressure ( ) is an extrapolated value that conceptually corresponds to the pressure to cause a doubling of the initial volume of the probe ( ), or Δ = . In estimating the extrapolated , two types of volumetric strain are defined for the PMT: (i) Δ ⁄ based on the initial volume, ; and (ii) Δ ⁄ based on the current volume, = + Δ , where Δ = measured volume change starting from beginning of inflation. There are five basic types of pressuremeters: • Prebored Menard • Prebored monocell • Push-in • Self-boring • Full-displacement types 5.2.3.1 Prebored Menard Pressuremeter The prebored pressuremeter is the most common and includes the original Menard type of pressuremeters. The Menard version includes three air-inflated guard cells: two situated above and one below the water-filled center testing probe cell. The testing time is slow because pressures for all three cells must be recorded, and readings are taken at predefined times before advancing to the next pressure increment. 5.2.3.2 Prebored Monocell Pressuremeter More recently, a simpler monocell probe has found favor in United States practice for prebored pressuremeters. This design omits the upper and lower guard cells and has only the middle probe cell that is water-inflated (Briaud 1992). A screw pump is convenient because each full revolution of the handle corresponds to a set injection volume of fluid, and thus ΔV is proportional to the number of rotations. Several commercial types of monocell probes are available (e.g., Texam, Oyo, and Pencel). 5.2.3.3 Self-Boring Pressuremeter Several versions of self-boring pressuremeters were simultaneously developed in England (Camkometer), France (PAFSOR), and Japan (Oyo). In these designs, the self-boring pressuremeter probe is lowered to the bottom of a borehole and then a small rotary cutter is used to insert the probe to the desired test depth. The probe is hollow, thus allowing the soil cuttings to be flushed out during advancement. Newer designs provide a water-jetting feature to cut soil at the front end and to facilitate soil removal.

78 Direct-Push In Situ Test Methods for Soils Several statically pushed in situ testing devices have been developed to facilitate direct measurements in soils and expedite field testing times. These direct-push methods often employ large (22.5 tons) hydraulic pushing systems that use either deadweight reaction or an anchoring system to force the probes into the ground. The hydraulic frames are mounted on trucks, tracks, and portable setups. The two most common direct-push in situ test methods are the CPT and DMT. 5.3.1 Cone-Penetration Testing The CPT is typically used to estimate the shear strength parameters, stiffness, coefficient of consolidation, and hydraulic conductivity of soils. The basic components of the CPT are illustrated in Figure 5-11. In this test, 25-ton hydraulic pushing systems mounted on wheeled trucks or track vehicles providing deadweight reaction are typically used to push the penetrometer into the ground. Also, anchored rigs can develop the necessary reaction forces for penetration. The standard CPT penetrometer has a cross-section area of 10-cm2 (1.55 in.2), although it is also common to use a cone with a cross-section area of 15-cm2 (2.33 in.2). The CPT penetrometer contains load cells, pressure transducers, inclinometers, or other sensor devices that measure the response of the ground as the penetrometer is advanced at a constant rate of 20 millimeters per second (mm/s) (0.8 in./s). The basic CPT test is conducted in accordance with ASTM D5778. In addition to the standard sizes, there are versions that include sizes between 1-cm2 (0.15 in.2) and 40- cm2 (6.2 in.2) in cross-section area for special applications (Mayne 2010, 2012). Smaller sizes are used for profiling in varved clays and highly stratified layers, as well as for soils requiring frequent pore pressure dissipation testing. Larger size penetrometers have found use in testing of gravelly soils. Table 5-3 lists the several types of cone penetrometers available and their applications.

79 Source: Paul Mayne Figure 5-11. Basic setup and equipment for (electric) CPT Table 5-3. Types of cone penetrometers Type of CPT Acronym Measurements Taken Applications Mechanical Cone Penetration MCPT qc (or qc and fs) on 8-in. (20-cm) intervals. Uses inner and outer rods to convey loads uphole Stratigraphic profiling, fill control, natural sands, hard ground Electric Friction Cone ECPT qc and fs (taken at 0.4- to 2-in. [1- to 5-cm] intervals) Fill placement, natural sands, soils above the groundwater table Piezocone Penetration Test CPTu and PCPT qc, fs, and either face u1, or shoulder u2 (taken at 0.4- to 2-in. [1- to 5-cm] intervals) All soil types (note: requires u2 for correction of qc to qt Piezocone with Dissipation CPTu Same as CPTu with monitoring of u1 or u2 during decay with time Normally conducted to 50% dissipation in silts and clays

80 Type of CPT Acronym Measurements Taken Applications Seismic Piezocone Test SCPTu Same as CPTu with downhole shear waves (Vs) at 3.3-ft (1-m) intervals Provides fundamental soil stiffness with depth: Gmax = ρt Vs2 Resistivity Piezocone Test RCPTu Same as CPTu with electrical conductivity or resistivity readings Detect freshwater–salt water interface. Index to contaminant plumes 5.3.1.1 Piezocone Penetration Testing The piezocone penetrometer is a hybrid device that marries an electric or electronic cone penetrometer together with a pore pressure probe. This allows the simultaneous measurement of dynamic pore pressures with cone tip resistance and sleeve friction. Porous filter elements can be positioned at various elevations on the penetrometer, including the apex and midface of the cone (Type 1), shoulder (Type 2), behind the sleeve (Type 3), as well as higher elevations (Lunne et al. 1997). The common configurations are shown in Figure 5-12. The standard position is the Type 2; however, there can be advantages to using Type 1 piezocone in desiccated clays for better profiling results and at sites requiring many dissipation tests for soil hydraulic conductivity evaluations. In soils with high fines content, a piezocone-type penetrometer is essential because the measured cone tip resistance ( ) is affected by pore pressures and must be corrected to the total cone tip resistance ( ). In dry soils there will be no pore pressures, and in clean sands the pore pressures will be approximately hydrostatic ( ≈ ), so that the value of ( ≈ ), so using a piezocone is not as critical. However, in many investigations, the subsurface conditions are not known, and thus a piezocone should be used so that the proper data and results are obtained. Source: Paul Mayne Figure 5-12. Common piezocone configurations for pore pressure readings

81 Example results for a CPTu are presented in Figure 5-13. (a) (b) (c) Source: Minnesota DOT Figure 5-13. Example results from CPTu showing (a) total cone resistance, qt, (b) sleeve friction, fs, and (c) pore pressures, u2 The piezocone is an efficient test for site characterization because it collects three independent and continuous readings related to soil behavior with depth. As a rule, the total cone tip resistance ( ) can be visually examined to get a basic understanding of the various layers and soil types encountered. Generally, if qt > 50 tsf (5 megapascals [MPa]), the soils are coarse-grained (i.e., sands), whereas if < 50 tsf (5 MPa), the soils are fine-grained (i.e., clay or silt). For the CPTu sounding shown in Figure 5-13(a), the profile indicates that most of the soil profile is sandy, yet distinct clay layers are found at depths of 3.3 ft (1 m), 11.5 ft (3.5 m), 23 to 39 ft (7 to 12 m), 49 ft (15 m), and 72 to 88 ft (22 to 27 m). Below the groundwater table, the pore pressure can also be used to assess soil type. The general rules for pore pressures at the shoulder ( ) are as follows: 1. Sandy soils exhibit ≈ 2. Intact clays and silts exhibit > 3. Fissured clays exhibit < For the example sounding in Figure 5-13(c) at depths below the water table at 29.5 ft (9 m), in the three sandy layers from 39 to 52 ft (12 to 16 m), 59 to 72 ft (18 to 22 m), and 88 to 118 ft (27 to 36 m), the measured readings are essentially hydrostatic (i.e., ). Intact clay layers show greater than hydrostatic. Furthermore, the ratio ∗ = ⁄ can provide an approximate idea about the clay consistency: soft: ∗ ≈ 2; firm: ∗ ≈ 4; stiff: ∗ ≈ 8, and hard: ∗ ≈ 15+. For stiff, fissured, overconsolidated clays, often < 0. As a rule, for onshore CPTu the maximum possible negative value of is approximately -1 atm. For the sleeve resistance ( ), the soil type is normally inferred from the friction ratio (FR): Rf = FR = 100∙ ⁄ (in percent). In the case of clean, uncemented sands, usually FR is less than 1 percent, while in MnDOT Wakota Bridge - CPTu C107 0 10 20 30 40 50 60 70 80 90 100 110 120 0 100 200 300 400 De pt h (fe et ) Cone Tip Resistance qt (tsf) 0 10 20 30 40 50 60 70 80 90 100 110 120 0 1 2 3 4 5 6 Sleeve Friction fs (tsf) 0 10 20 30 40 50 60 70 80 90 100 110 120 0 1 2 3 4 5 6 7 8 9 10 Porewater Pressure u2 (tsf) INTACT CLAY: u2 > u0 SAND: u2 ≈ u0 qt u2 fs Groundwater at 30 ft CLAY SAND 50 tsf (5 MPa)

82 most clays FR is greater than 4 percent. In highly organic clays and peats, FR is greater than 6 percent. In very sensitive clays, however, FR may be less than 1 percent, similar to sands; thus, FR by itself should not be considered reliable for identifying soil type. 5.3.1.2 Piezo-dissipation Testing When an advancing penetrometer is halted, the induced pore pressures will begin to dissipate and seek equilibrium conditions that reflect hydrostatic conditions ( ). By measuring the rate of decay of Δ with time, the data can be used to evaluate the coefficient of consolidation ( or ) and the hydraulic conductivity (K) of the soil. Often, the piezo-dissipation data are plotted with pore pressure vs. log or square root time, similar to laboratory consolidation tests. Two primary types of response are usually observed: monotonic and dilatory. The simplest response is a monotonic-type dissipation whereby the measured pore pressures are highest during penetration, and then during dissipation, the readings decrease with time, eventually reaching u0. A dilatory response may also be observed, often in overconsolidated or fissured soils, where the pore pressures first rise with time to a peak value, then decrease subsequently to reach u0. Figure 5-14 shows a series of piezo-dissipation measurements from a highway embankment in North Carolina. Monotonic-type responses were observed at depths of 13.8 and 25 ft (4 and 7.6 m), while dilatory curves were obtained at depths of 18 and 27.2 ft (5.4 and 8 m). In many clays and soils of low hydraulic conductivity, it is inconvenient to wait for 100 percent dissipation to to occur because of the very long times required for consolidation. For practicality, a convenient reference mark is taken at which is the time required to reach an average degree of consolidation equal to 50 percent. Source: Paul Mayne (based on data from North Carolina DOT) Figure 5-14. Example piezo-dissipation tests from CPTu 5.3.1.3 Seismic Piezocone Test The seismic piezocone is a hybrid device that incorporates one or more geophones in the penetrometer to permit seismic downhole test (DHT) and the measurement of shear wave velocity with depth in addition

83 to penetration resistances. The general setup and procedures of the test (ASTM D5778 and D7400) are depicted in Figure 5-15. The CPT/CPTu is conducted over a 3.3-ft (1-m) stroke interval and then stopped to allow DHT. Then, after DHT, the CPT/CPTu portion is resumed. The SCPTu is very versatile because it obtains information about the stratigraphy, strength, stiffness, stress state, and hydraulic conductivity of the ground with up to five separate measurements: , , , , and shear wave velocity ( ). In the simplest arrangement, a single horizontal geophone (i.e., velocity transducer) is positioned above the friction sleeve and a pseudo-interval test is conducted using a surface source with its axis parallel to the geophone. A biaxial geophone (two orthogonal horizontal geophones at same elevation) offers an improved arrangement in case the penetrometer is inadvertently rotated during penetration. A superior setup for the seismic portion is called true-interval DHT, where two geophones are used behind the penetrometer usually at 3.3-ft (1-m) vertical spacing, so that the same seismic wavelet passes by the upper and lower geophones. If proper care and attention to detail are given, all the above methods essentially give the same results. Source: Paul Mayne Figure 5-15. SCPTu The generation of seismic wavelets is facilitated by using an automatic shear wave source (i.e., autoseis). This helps to provide repeatable waveforms so the delineation of shear wave arrivals is improved via cross- correlation, rather than the first arrival time or crossover points that are identified using paired sets of right and left strikes. Autoseis units have been developed that rely on paired electric solenoids, hydraulic, pneumatic, and electromechanical impact type wave generation, as well as steady-state sources and heavy mass vibrators. An illustrative example of SCPTu results are presented in Figure 5-16. Beneath shallow sediments of the Norfolk Formation, the presence of a stiff Miocene-age clayey calcareous sand (Yorktown Formation) is evident at depths below 30 ft (10 m) where the pore pressure readings go well below hydrostatic values. Also, the profile of . starts with very low values of less than 500 ft/s (152 meters per second [m/s]) in the shallower and weaker Holocene soils of the Norfolk Formation and increases to 1,000 ft/s (300 m/s) in the lower and stronger Yorktown unit.

84 Source: ConeTec Figure 5-16. Example SCPTu soundings in sediment Continuous-interval shear wave measurements can be obtained with Vs readings as frequent as the cone penetrometer values, on the order of several centimeters (Ku et al. 2013). An example continuous-interval SCPTu sounding is shown in Figure 5-17. Beneath an approximate 3.3-ft (1-m) fill layer, there is a silty layer to 23 ft (7 m), underlain by sands to 98 ft (30 m) which overlie soft silty clays until the termination depth of the sounding at 148 ft (45 m). Also shown are the Vs results from conventional DHT at 3.3-ft (1- m) intervals in a separate benchmark sounding to verify the approach. Very good agreement is seen with both procedures where the value of Vs increases from 295 ft/s (90 m/s) at shallow depths to about 820 ft/s (250 m/s) at the final depths of the sounding at 148 ft (45 m).

85 Source: Mayne and Woeller (2015) Figure 5-17. Example continuous-interval SCPTu 5.3.2 Flat Plate Dilatometer Test The DMT is primarily used to estimate the stiffness of the soil and the lateral stress coefficient. The DMT is conducted according to ASTM D6635. The testing procedures consist of hydraulically pushing a high- strength stainless steel blade into the ground at 0.66-ft (0.2-m) increments. At each depth increment, a flexible circular steel membrane (diameter of 60 mm) is inflated horizontally into the soil using nitrogen, and two readings of pressure are taken. The first reading (A) is collected in about 15 seconds when the membrane is flush with the flat blade (δ = 0), and the second reading (B) taken at δ = 1.1 mm, approximately 30 seconds later. As soon as the B reading is obtained, the membrane is quickly deflated, and the blade is pushed to the next test depth. The components of the DMT are illustrated in Figure 5-18.

86 Source: Paul Mayne Figure 5-18. DMT procedures and measurement readings Once the DMT data is processed, the field-measured A and B readings are corrected for membrane stiffness in air and renamed: p0 = contact pressure and p1 = expansion pressure (see Fig. 5-18). These two pressures are used to calculate three DMT indices: 1. = ( − ) ( − )⁄ = material index which serves to identify soil type, (i.e., clay when ID < 0.6 and sand when ID > 1.8). 2. = 34.7( − ) = (1 − )⁄ = dilatometer modulus that provides a measure of soil stiffness. 3. = ( − )⁄ = horizontal stress index that is used to profile soil strength and stress history. Example and pressure readings with depth from a DMT sounding performed in southeastern Virginia are presented in Figure 5-19. The corresponding profiles of ID, ED, and KD are shown in Figure 5-20. The material index (ID) clearly indicates the occurrence of shallow interbedded sands and silts with a transition at a depth 18 ft (5.5 m) to a silty clay layer that extends to 56 ft (17 m), underlain by silty sands through to the termination depth of the sounding at 100 ft (30 m).

87 Source: ConeTec Figure 5-19. Example DMT measured pressure profiles Source: ConeTec Figure 5-20. Example DMT index profiles Some additional options for the DMT include (i) force measurements during blade penetration, (ii) A- dissipation readings, and (iii) C readings. Force readings during penetration of the blade are discussed by

88 Campanella and Robertson (1991). The forces can be used to evaluate the blade resistance ( ) that is a stress measured during penetration. At selected depths, DMT dissipation testing can be conducted by recording the decay of A readings with time (Failmezger and Anderson 2006). The dissipation procedures for DMT are accomplished by repeatedly inflating the membrane to the A position, then releasing the pressure (i.e., no B reading is obtained). The time corresponding to a change in slope in a plot of vs. logarithm of time is termed , which is used in evaluating the coefficient of consolidation, . A third pressure reading (C) can be taken at δ = 0 during the membrane deflation after the B reading is obtained. It is similar to the A reading but is measured during deflation rather than inflation. The C reading is corrected as per the A reading and called . In sands, corresponds to the hydrostatic pore pressure ( ). In Situ Tests for Rock Several in situ tests are available for the field testing of intact and fractured rocks. The following are the most common in situ tests for rock: • PLT • FJT • Rock dilatometer • Large field DS test • Rock borehole shear test • Borehole cameras 5.4.1 Plate Load Tests on Rock PLT on rock is a common in situ testing method for evaluating the rock mass stiffness (Bieniawski 1981). A PLT on rock is similar to a PLT on soil, except that much larger plates, reaction system, and hydraulic forces are required for tests on rock. The testing procedure consist of applying stress using a jacking system and measuring the resulting displacements. The stress can be applied either vertically or horizontally depending on the nature of the proposed construction. For example, investigation studies for shallow and deep foundations require vertical PLTs, and investigations involving tunnels require horizontal PLTs. In addition to measuring the plate displacement, borehole extensometers can also be installed on large projects so that displacements at different depths inside the rock mass can also be measured during loading. Test procedures for PLT on rock follow ASTM D4394 for rigid plates and D4395 for flexible plates. The procedures are similar to those for PLT on soils per AASHTO T221 and T222. The test results are presented in terms of stress vs. displacements graphs (Figure 5-21). Generally, the results are analyzed using well-known elasticity solutions to back calculate an equivalent modulus ( ), as discussed by Hoek (2007).

89 Source: data from Kavur et al. (2015) Figure 5-21. Example PLT on limestone 5.4.2 Flat Jack Tests FJTs are used for two purposes in rock mass characterization: (i) evaluate the geostatic in situ state of stress and (ii) determine rock mass stiffness (Bieniawski 1981, Deklotz and Boisen 1970). For geostatic in situ state of stress determination, several paired sets of pins are installed at known distances on the rock face (Figure 5-22). A slot no more than 3 in. (7.6 cm) wide with a length-to-width ratio greater than five is cut between the paired pins. A square flat jack at least 2 ft (0.6 m) wide is installed and grouted into the slot and later pressurized after the grout has cured (Rocha 1970). The pressure required to return the pins to their original positions is interpreted as the geostatic stress state ( ) perpendicular to the direction of the slot opening. Preferably three separate FJTs using orthogonal slots should be cut in the rock surface to fully determine the in situ state of stress: , , and . In addition to measuring the in situ stress, the FJT can be further pressurized and displacement measurements taken to assess the elastic modulus (E') of the rock material via the theory of elasticity.

90 Source: Paul Mayne Figure 5-22. Cross section of flat jack setup per ASTM D4729 5.4.3 Rock Dilatometer A rock dilatometer test is used to evaluate rock mass stiffness. The testing procedure involves placing a long cylindrical probe into a rock-cored borehole, inflating the probe laterally, and measuring the resulting displacements (Rocha 1970). A special directional rock dilatometer has been developed to permit evaluations of stress anisotropy. 5.4.4 Large Field Direct Shear Test Large-scale DS tests can be conducted to measure the shear strength of natural rock outcrops in surficial exposures or in excavations. These tests are analogous to laboratory DS tests where a normal load is first applied vertically, then a horizontal load is used to create a shearing stress across the rock material. The large in situ DS test can also be used along joints or shear planes in the rock to measure the peak and residual strength of the discontinuities (Goodman 1970). Test equipment and procedures for large in situ DS test of rock joints are provided in ASTM D4554. 5.4.5 Rock Borehole Shear Test The (Iowa) borehole shear test, originally developed for soils, was modified to allow testing in rocks (Yang et al. 2006) and called the rock borehole shear test. The rock borehole shear test is used to measure the shear strength of rock. A downhole probe with paired shear plates is lowered to the desired test elevation. Pressure is applied to the plates to obtain the normal stress, and a tether is pulled to create a shear force on the plates. The derived shear stress (τ) vs. normal stress (σ') directly provides the Mohr-Coulomb strength envelope as shown in Figure 5-23. = + tan where = effective cohesion intercept = effective friction angle Rock Face Slot cut into rock for test Rock Face Grout Flatjack inserted into slot Pump Pressure GaugeLimestone Rock Formation RQD = 75 RMR ≈ 55

91 Source: Dick Handy, Handy Geotechnical Equipment Figure 5-23. Rock borehole shear test in sandstone 5.4.6 Measurements for Rock Discontinuities Optical and acoustic televiewers provide a continuous, 360° view of the borehole wall that allows rock mass discontinuities to be identified and characterized. Both devices can be oriented within the borehole so that the absolute orientation of features (e.g., bedding planes) can be measured. The OTV uses visible light whereas the ATV uses ultrasound to image the borehole wall. OTV and ATV surveys are used to provide information regarding locations and orientations of joints, bedding planes, faults, shears, and other naturally occurring rock mass discontinuities. A more detailed summary of optical and acoustic televiewers is provided in Section 4.4.2.4.

92 Chapter 5 References Batchelor, C., G. Goble, J. Berger, and R. Miner. 1995. Standard Penetration Test Energy Measurements on The Seattle ASCE Field Testing Program. GRL Report to American Society of Civil Engineers (ASCE) and Univ. of Washington. Bieniawski, Z.T. 1981. "Experience with In Situ Measurement of Rock Deformability." Geophysical Research Letters, Vol 8, No. 7, pp. 675–677. Boulanger, R.W., and I.M. Idriss. 2014. CPT and SPT Based Liquefaction Triggering Procedures. Report No. UCD/CGM- 14/01, Center for Geotechnical Modeling, University of California-Davis. Briaud, J-L. 1992. The Pressuremeter. Swets and Zeitlinger, Lisse, The Netherlands. Broms, B.B., and N. Flodin. 1988. "History of Soil Penetration Testing." Penetration Testing, Vol. 1 (Proc. ISOPT-1, Orlando) Balkema, Rotterdam. pp. 157–220. Campanella, R.G., and P.K. Robertson. 1991. "Use and Interpretation of a Research Dilatometer." Canadian Geotechnical Journal, Vol. 28, No. 1, pp. 113–136. Chandler R. J. 1988. "The In Situ Measurement of the Undrained Shear Strength of Clays Using the Field Vane." Vane Shear Strength Testing in Soils: Field and Laboratory Studies, ASTM Special Tech. Publ.1014. West Conshohocken, Pennsylvania. pp. 13–44. Davidson, J.L., J.P. Maultsby, and K.B. Spoor. 1999. Standard Penetration Test Energy Calibrations. Report WPI 0510859 by University of Florida - Gainesville to Florida Dept. Transportation, Tallahassee. Deklotz, E.J., and B.P. Boisen. 1970. "Development of Equipment for Determining Deformation Modulus and In-Situ Stress by Means of Large Flatjacks." Determination of the In-Situ Modulus of Deformation of Rock Masses, Special Technical Publication 477, ASTM, West Conshohocken, Pennsylvania. pp. 117–125. Failmezger, R.A., and J.B. Anderson (eds). 2006. Proceedings 2nd International Conference on Flat Dilatometer. Arlington, Virginia. Goodman, R.E. 1970. "The Deformability of Joints." Determination of the In Situ Modulus of Deformation of Rock Masses, Special Tech. Publ. 477, ASTM, West Conshohocken, Pennsylvania. pp. 174–196. Hoek, E. 2007. Practical Rock Engineering. RocScience Inc., Toronto, ON. Honeycutt, J.N., S.E. Kiser, and J.B. Anderson. 2014. "Database Evaluation of Energy Transfer for CME Automatic Hammer SPT." Journal of Geotechnical & Geoenvironmental Engineering, Vol. 140, No. 1, pp. 194–200. Kavur, B., N. Stambuk, and P. Hrzenjak. 2015. "Comparison between Plate Jacking and Large Flat Jack Test Results of Rock Mass Deformation Modulus." International Journal of Rock Mechanics and Mining Sciences, Vol. 73, pp. 102– 114. Kelley, S.P., and J.E. Lens. 2010. Evaluation of SPT Hammer Energy Variability, Windsor, VT. Report No. 2010-3/Contract 0984735 GeoDesign Inc. Montpelier, Vermont. Ku, T., P.W. Mayne, and E. Cargill. 2013. “Continuous-Interval Shear Wave Velocity Profiling by Auto-Source and Seismic Piezocone Tests.” Canadian Geotechnical Journal, Vol. 50, No. 1, pp. 382–390. Kulhawy, F.H., and P.W. Mayne. 1990. Manual for Estimating Soil Properties for Foundation Design. Report EL-6800, Electric Power Research Institute, Palo Alto. Lunne, T., P.K. Robertson, and J.J.M. Powell. 1997. Cone Penetration in Geotechnical Practice. EF Spon/Blackie Academic-Routledge, London. Mayne, P.W. 2010. "Regional Report for North America." Proceedings, 2nd International. Symposium on Cone Penetration Testing (CPT'10). Vol. 1, Huntington Beach, California. pp. 275–312. Mayne, P.W. 2012. SOA Report: Geotechnical Site Characterization in the Year 2012 and Beyond. State-of-the-Art and Practice in Geotechnical Engineering. GSP 226, GeoCongress, Oakland, CA, ASCE Press, Reston, Virginia. Mayne, P.W. and Woeller, D.J. 2015. “Advances in Seismic Piezocone Testing.” Proc. XVI European Conference on Soil Mechanics & Geotechnical Engineering, Vol. 6, Edinburgh, United Kingdom. pp. 3005–3009. Peuchen, J., and P.W. Mayne. 2007. “Rate Effects in Vane Shear Testing.” Proceedings of the 6th International Offshore Site Investigation and Geotechnics Conference: Confronting New Challenges & Sharing Knowledge, Society for Underway Technology, United Kingdom. pp. 259–266.

93 Robertson, P.K. 1986. "In Situ Testing and its Application to Foundation Engineering." Canadian Geotechnical Journal, Vol. 23, No. 6, pp. 573–594. Rocha, M. 1970. "New Techniques in Deformability Testing of In-Situ Rock Masses." Determination of the In-Situ Modulus of Deformation of Rock Masses, Special Tech. Publication 477, ASTM, West Conshohocken, Pennsylvania. pp. 39–57. Seed, H.B., K. Tokimatsu, L.F. Harder, and R.M. Chung. 1985. “The Influence of SPT Procedures in Soil Liquefaction Resistance Evaluations.” Journal of Geotechnical Engineering, Vol. 111, No. 12, pp. 1425–1445. Sjoblom, D., J. Bischoff, and K. Cox. 2002. "SPT Energy Measurements with the PDA." Proceedings of the 2nd International Conference on the Application of Geophysical and NDT Methodologies to Transportation Facilities & Infrastructure, Conference sponsored by FHWA, TRB, and CALTRANS. Skempton, A.W. 1986. “Standard Penetration Test Procedures and Effects in Sands of Overburden Pressure, Relative Density, Particle Size, Ageing, and Over Consolidation.” Geotechnique, Vol. 36, No. 3, pp. 425–447. Webster, S.D., and M. Rawlings. 2009. Standard Penetration Test Energy Measurements. GRL Engineers Inc. Job. No. 099015 Report to AECOM, Vernon Hills, Illinois. Yang, H., D.J. White, and V.R. Schaefer. 2006. “In Situ Borehole Shear Test and Rock Borehole Shear Test for Slope Investigation.” Site and Geomaterial Characterization (GSP 149, GeoShanghai), pp. 293–298. Youd, T.L., M. Idriss, R.D. Andrus, I. Arango, G. Castro, J.T. Christian, R. Dobry, W.D.L. Finn, L.F. Harder, Jr., M.E. Hynes, K. Ishihara, J.P. Koester, S.S.C. Liao, W.F. Marcuson, III, G.R. Martin, J.K. Mitchell, Y. Moriwaki, M.S. Power, P.K. Robertson, R.B. Seed, and K.H. Stokoe, II. 2001. “Liquefaction Resistance of Soils: Summary Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction Resistance of Soils.” Journal of Geotechnical and Geoenvironmental Engineering, Vol. 127, No. 10, pp. 817–833.