Cast-in-Place Concrete Connections for Precast Deck Systems (2011)

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2 formwork such as steel or concrete panels, place deck reinforcement, cast deck concrete, and remove formwork if necessary. This project focused on systems that reduce the need to place and remove formwork thus accelerating on-site construction and improving safety. The three systems considered in NCHRP 10-71 to accomplish these objectives were identified during the 2004 Prefabricated Bridge Elements and Systems International Scanning tour (International Scanning Study Team (2005)). The scanning tour visited France, Belgium, Japan, and the Netherlands with eleven participants representing FHWA, State Departments of Transportation, National Association of County Engineers, industry, and academia. The study team developed a series of recommendations related to prefabricated elements and systems to be used for superstructure systems, along with substructure systems and movement systems for rapid replacement and construction. Examples of these systems include a French precast slab superstructure (Poutre Dalle) with overlapping looped reinforcement that extends into the longitudinal CIP connections, a Japanese full-depth deck panel system with similar looped reinforcement in the transverse CIP connections, and a Japanese full-depth deck flange system with similar looped reinforcement in the longitudinal CIP connection. Additionally, other deck connection details were identified for development through this research. Three of the superstructure systems were specifically addressed in this project. These systems included: (1) a precast composite slab span system (PCSSS) for short to moderate span structures based on the French Poutre Dalle system, (2) full-depth prefabricated concrete decks, and (3) deck joint closure details (e.g., decked-bulb-tee (DBT) flange connections) for precast prestressed concrete girder systems for long span structures. Each system uses precast elements that are brought to the construction site ready to be set in place and quickly joined together. Depending on the system, the connections are either transverse (i.e., across the width of the bridge) or longitudinal (i.e., along the length of the bridge). The first system, PCSSS, is an entire bridge system; whereas the other two systems investigated in the project represented transverse and longitudinal joint details to transfer moment and shear in precast deck panels and flanges of decked bulb tees. Because of the similarities in the latter two types of systems, they are grouped together in this report. Two types of connection concepts were explored with these details, looped bar details and two layers of headed bar details. Although both types of systems performed adequately in initial tests, the looped bar systems were deemed to be more practical for construction purposes and were investigated in the subsequent tests. To implement these promising new systems, design guidelines and standard details were established. The developed guidelines also address the durability of the connection concepts including crack control. The PCSSS bridge design guidelines cover both component and system issues including “spalling” reinforcement, load distribution, effect of restraint moments, composite action, and reinforcement to control reflective cracking. Critical goals for the precast panel and decked-bulb-tee flange connections included minimizing the required connection depth to accommodate the loop bar while ensuring concrete cover, achieving durable material properties for the CIP concrete, and preventing moisture from penetrating the interface between the precast and CIP concrete. The research team included individuals with expertise in structural analysis and testing, materials testing, bridge design, precast concrete fabrication, bridge construction, specification development, and examples and tools development. The researchers at the University of Minnesota had been involved with the first PCSSS bridge implementations in the State of Minnesota, for which they instrumented one of the first two bridges constructed for the Minnesota Department of Transportation (Mn/DOT). In addition, they conducted an investigation on a two-span PCSSS laboratory bridge specimen to investigate a number of variations with the system. Following that study, Mn/DOT agreed to make the two-span bridge available for testing in association with the NCHRP 10-71 project.

3 The following sections summarize the objective and list of tasks associated with the NCHRP 10-71 project, Cast-in-Place Concrete Connections for Precast Deck Systems. 1.1. Scope of Study The objective of this project was to develop guidance for the design and construction of durable CIP reinforced concrete connections for precast deck systems that emulate monolithic construction. This was accomplished through the following ten tasks. 1.1.1. Task 1 – Review relevant practice, performance, data, and research findings Reviewed relevant practice, performance data, research findings, physical test results, and other information related to the design, fabrication, and installation of CIP reinforced concrete connections for precast deck systems that emulate monolithic construction. This information was gathered from technical literature and from unpublished experiences of engineers, bridge owners, and others. This review included the looped reinforcement details identified in Japan and France during the 2004 Prefabricated Bridge Elements and Systems International Scan. 1.1.2. Task 2 – Develop detailed design, fabrication, construction, and performance criteria Developed detailed design, fabrication, construction, and performance criteria that were selected to provide durability, strength, fatigue resistance and rapid construction. 1.1.3. Task 3 – Develop conceptual designs for CIP reinforced concrete connections Developed several conceptual designs for CIP reinforced concrete connections, including a comprehensive longitudinal connection detail for precast slab superstructures and three connection concepts for longitudinal or transverse connections between full-depth deck panels or deck flanges. The connection concepts were developed to not require overlays or post-tensioning. Emphasis was placed on increasing construction speed while achieving durability and ride quality. 1.1.4. Task 4 – Develop an updated and detailed work plan Developed an updated and detailed work plan for numerical and experimental evaluation of the conceptual designs based on the results of Tasks 2 and 3, including the development of large-scale connection test specimens. 1.1.5. Task 5 – Submit an interim report The interim report was completed and submitted to seek approval to move forward with the work plan.

4 1.1.6. Task 6 – Execute the approved work plan for evaluation of the connections This task represented the bulk of the effort for this project during which the approved work plan was executed for the evaluation of the connections. 1.1.7. Task 7 – Prepare a connection design, detailing guide, and construction guide Prepared a comprehensive connection design and detailing guide, including a listing of all design steps with examples as needed for the connections that met the criteria established in Task 2. 1.1.8. Task 8 – Develop specification language and commentary Developed specification language and commentary for recommended changes to the AASHTO LRFD Bridge Design Specifications and the AASHTO LRFD Bridge Construction Specifications as necessary to implement the recommended connection details. 1.1.9. Task 9 – Submit the products of Tasks 7 and 8 and the Draft Final Report The products of Tasks 7 and 8 were submitted for panel review and comment. 1.1.10. Task 10 – Final Report The research study has been documented in this final report. It includes the design and detailing guide with recommended modifications to the design specifications and a list of design steps and examples. 1.2. Introduction to Precast Composite Slab Span Systems (PCSSS) Precast composite slab span systems (PCSSS) are a promising technology for the implementation of rapid construction techniques for bridge construction. The bridge systems are composed of precast, inverted-T sections, fabricated off-site and delivered to the jobsite ready for erection. The inverted-T sections are assembled such that no formwork is required prior to the placement of the CIP deck, which considerably reduces construction time related to the placement and removal of formwork. Transverse load transfer is achieved through the development of transversely oriented reinforcement protruding from the precast members. Furthermore, improved quality of the main superstructure can be achieved due to the rigid quality control associated with the fabrication of precast members, which may be difficult to achieve in cast-in-place (CIP) bridge construction. The current study included an investigation of the literature related to precast composite slab span bridge systems and other relevant topics valuable in the development of PCSSS bridges. Included in this review was the work completed during a study commissioned by Mn/DOT (Smith et al., 2008) regarding a field and laboratory implementation of a PCSSS bridge. The laboratory bridge specimen utilized during the Mn/DOT study was subsequently made available for use with the NCHRP 10-71 project. The research conducted during the NCHRP 10-71 study included numerical parametric studies and further experimental studies of variables deemed to be important from the Mn/DOT study including the effects of reinforcement to control spalling, connection details for crack control, cyclic loading effects on crack development, and composite action between the precast and CIP.

5 Because of its similarity to slab-span bridge systems, the applicability of the AASHTO design provisions for slab-span systems were investigated in the study. The two primary considerations that distinguish PCSSS bridges from slab-span bridges are (1) the required reinforcement to control reflective cracking above the longitudinal joint between the precast flanges, and (2) the effect of restraint moments due to the composite nature of the system. With regard to the issue of reflective crack control, in addition to a numerical investigation regarding the effect of the transverse reinforcement, the issue was also studied in laboratory investigations of two large-scale laboratory specimens (i.e., Concept 1 and 2 bridges), as well as in subassemblage test specimens specifically designed to investigate crack control. The effect of restraint moments in the PCSSS were investigated numerically and experimentally in a previous study by the researchers, the results of which can be found in Smith et al. (2008) and Eriksson (2008). Other considerations for the design of the PCSSS investigated in the study included the composite action of the precast and CIP which was investigated through ultimate load tests of the Concept 1 and 2 bridges and the spalling reinforcement detail of the precast section. Current design requirements to control the end zone stresses in prestressed concrete members were developed for I-sections and are not directly applicable to the PCSSS precast inverted-T sections. The current design recommendations were found to be conservative for shallow inverted-T sections and unconservative for deeper inverted-T sections. There were a few considerations not included in the laboratory research or numerical study such as the connection between the precast elements and the substructure. These details were investigated primarily by means of examination of structural plans for existing PCSSS structures. 1.3. Introduction to Longitudinal and Transverse Joints in Decked Bulb-T (DBT) and Full- Depth Precast Panel on Girder Systems Two issues that limited the PCSSS bridge concept with regard to the potential for accelerated bridge construction applications were (1) the significant use of CIP to complete the composite system, which would slow the construction process, and (2) the limitation of the system to short- to moderate- span lengths. As a consequence, NCHRP 10-71 included the study of CIP connection concepts that minimized the use of CIP by limiting its application to the joints between the flanges of decked bulb-T’s or between full-depth precast deck panels on girders. The investigated joints used two layers of reinforcement to provide the ability to transfer moment as well as shear through the deck. Two types of details were investigated to reduce the width of the joint: U-bar details (with deformed wire reinforcement (DWR) and stainless steel (SS)) and headed reinforcement details. Tests were included to evaluate the effectiveness of the different connection concepts, and the most promising connection concept in terms of behavior, constructability and cost was subjected to a series of parametric tests to further refine the joint details. The investigation included the development of performance specifications to achieve high performance durable closure pour (CP) materials for both overnight cure and 7-day cure applications. Numerical studies were conducted with a number of variations to investigate service static and fatigue loadings that might be expected in the longitudinal and transverse joint connection concepts. Large-scale longitudinal and transverse jointed specimens were fabricated to investigate the flexure and flexure-shear behavior of the longitudinal joints and the tension behavior of the transverse jointed specimens. The large-scale specimens were fabricated with the most promising connection detail which was a U-bar connection concept fabricated with deformed wire reinforcement (DWR). The specimens were subjected to static and fatigue tests with the loads determined in the numerical parametric study.

6 The tests were evaluated in terms of load-deformation response, strain distribution, crack control, and strength. 1.4. Organization of Report As mentioned in Section 1.0, this report covers two very different systems: (1) the precast composite slab-span system (PCSSS), which is an entire bridge system, and (2) transverse and longitudinal cast-in- place connection concepts to transfer moment and shear between precast deck panels and the flanges of precast decked bulb-Ts. The report is separated accordingly. The research contained in Chapters 2 through 7 includes the documentation associated with Tasks 1 through 8 related to the development of PCSSS bridges. Chapters 8 through 14 include the documentation associated with Tasks 1 through 8 related to the development of longitudinal and transverse connection concepts between full-depth deck panels and decked bulb-T flanges. The chapters associated with the PCSSS bridge system are organized as follows. Chapter 2 provides a literature review of the system and associated parameters of interest including literature on crack control, composite action of cast-in-place (CIP) and precast systems, and spalling design requirements. Chapter 3 provides background on the system which was initially implemented in the U.S. by the Minnesota Department of Transportation (Mn/DOT). The chapter contains the results of a field and laboratory investigation sponsored by Mn/DOT which included an investigation of live load distribution and the effects of restraint moments. Chapter 4 presents a summary of a parametric study to investigate the range of applicability of PCSSS bridges. Numerical studies are summarized in this chapter, which were conducted to investigate a number of parameters including transverse reinforcement spacing for crack control, applicability of slab-span design recommendations for PCSSS in terms of live load distribution factors and skew effects. The chapter also summarizes the numerical and experimental results associated with end zone stresses in the inverted-T precast sections used to fabricate the PCSSS, and a review of support conditions that have been used for the systems in the field. The chapter concludes with the results of a numerical study used to determine the magnitude of the patch load to be applied in the laboratory investigations of the PCSSS Concept 1 and Concept 2 bridges. Chapter 5 presents a summary and results of the tests on the Concept 1 and 2 laboratory bridges. Subassemblage tests used to investigate detailing of the joint region for crack control are summarized in Chapter 6. Conclusions of the PCSSS investigation are provided in Chapter 7. As mentioned above, the investigation of the longitudinal and transverse connection concepts between decked bulb-Ts or full-depth precast deck panels is summarized in Chapters 8 through 14. Chapter 8 contains a brief introduction. Rather than having a separate chapter on literature review for the longitudinal and transverse connection concepts, the literature is summarized in the respective chapters where appropriate. Chapter 9 summarizes the preliminary tests on monolithic specimens containing the U-bar (deformed wire reinforcement (DWR) and stainless steel (SS)) and headed reinforcement to determine the most promising connection detail; it presents the results of parametric tests used to further refine details of the connection concept. Chapter 10 detailed the numerical parametric study conducted to determine loads to be applied to large-scale longitudinal and transverse joint specimens described in later chapters. Chapter 11 details the development of performance specifications for overnight-cure and 7-day-cure closure pour (CP) materials considered for the joints. The chapter also presents the results of tests on a number of prospective CP materials used to determine the most promising CP candidate materials in each category (i.e., for overnight cure and 7-day cure) that were subsequently used in the fabrication of the large-scale longitudinal and transverse joint specimens. Chapters 12 and 13 summarize the static and fatigue loading tests on the large-scale longitudinal and

7 transverse joint specimens, respectively, and Chapter 14 presents a summary of the conclusions of the study. Appendix A contains the recommended design recommendations including suggested changes to the AASHTO LRFD Bridge Design code and commentary that were developed during the study, in accordance with Task 8. Appendix B contains detailed design examples in accordance with Task 7. The results of the phone survey conducted at the initial stages of the project to gather information including the respondents’ experience with similar systems, their input on important performance criteria, and their feedback on proposed connection concepts are contained in Appendix C. Appendix D contains information on the instrumentation of one of the early implementations of the PCSSS in the State of Minnesota. Appendices E and F contain the information on the instrument locations in the two laboratory PCSSS bridge specimens. The coring analysis of the PCSSS laboratory bridge and subassemblage test specimens are contained in Appendix G. The subassemblage sectional design calculations and information on the instrumentation locations are contained in Appendices H and I, respectively.

8 Chapter 2 PCSSS: Literature Review 2.0 Introduction to Literature Review The design and implementation of the experimental and numerical studies regarding precast composite slab span systems (PCSSS) associated with the NCHRP 10-71 project were completed after consideration of available previous research. Because the PCSSS was a relatively new concept, the existing literature associated with the system was limited. The system was based on the Poutre Dalle bridge system developed in France, but there were no published results regarding the performance of that system. The designs associated with the original implementation of the PCSSS in the U.S. were based on the AASHTO provisions for slab-span systems because of their many similarities. Major differences exist, however, between the PCSSS and slab-span bridge systems including the composite precast / cast-in-place (CIP) nature of the PCSSS which can result in the development of restraint moments in continuous bridges due to the different time-dependent properties of the component materials. In addition, the precast portion of the PCSSS had issues to be resolved regarding the applicability of AASHTO reinforcement provisions to control end cracking due to spalling stresses, and issues associated with the development of details to control reflective cracks that might develop above the interface between the precast flanges or the CIP-precast web interface. 2.1. Poutre Dalle System During the FHWA/AASHTO 2004 International Scan Tour, rapid construction techniques used for prefabricated bridges in Europe were investigated (Hagen, 2005). The Poutre Dalle system used in France showed promise for rapid construction of short to moderate span bridges, shown in Figure 2.1.1. The system was designed to provide longitudinal moment capacity through the longitudinally oriented precast beams made composite with a CIP deck surface. Transverse load distribution was achieved through the development of transversely oriented reinforcement protruding from the precast section in the CIP concrete placed in the trough region created by adjacent panels. The Minnesota Department of Transportation developed a precast composite slab span system (PCSSS), described in Chapter 3, that was based on the Poutre Dalle concept. Mn/DOT implemented the first two PCSSS bridges in 2005, both of which had been previously planned for CIP slab construction. One of them was instrumented to be monitored in two ways: to investigate the magnitude and location of reflective cracking between adjacent inverted-T sections and to examine the conitnuity over the piers. Additionally four Mn/DOT PCSSS secctions were delivered to the University of Minnesota structures laboratory for additional study for possible refinements to the design and detailing of the system, described in Chapter 4.

9 Figure 2.1.1: Photograph of precast section used in Poutre Dalle System (Hagen, 2005) 2.2. Crack Control Reinforcement, Frosch et al., 2006 Cracking in bridge systems must be adequately controlled for aesthetic as well as performance and durability considerations. The presence of cracking in bridge superstructures introduces a means of ingress for chlorides and other contaminants which can be detrimental to the performance of the embedded reinforcement. This concern is especially applicable in regions with harsh environments, such as those with freeze-thaw cycles, which are also usually coupled with the application of large amounts of salt to bridge superstructures. Crack control requires the use of adequately sized and spaced reinforcement. Frosch et al. (2006) provided an extensive investigation of cracking in bridge decks and developed associated design recommendations for reinforcement to control cracking. Although Frosch’s application was focused on the crack control of bridge decks, the findings of his research were applied to the NCHRP 10-71 study to investigate their applicability in the control of potential reflective cracking above the longitudinal joint between the precast flanges. Frosch et al. (2006) completed an experimental research program through the observation of four highway bridge decks in Indiana, three of which were instrumented by the researchers. The fourth bridge was instrumented in a previous research study by Radabaugh (2001), the primary objective of

10 which was to observe early age behavior of bridge decks exposed to field conditions. The bridges selected for the study included three two-span continuous bridges, with equal spans of 76, 101, and 123 ft. The fourth bridge was a five span bridge with spans of 39, 63, 77, 63, and 39 ft. The decks in the three two-span bridges were constructed with two layers of steel reinforcement for crack control, while the five span bridge was built with GFRP for the top layer of reinforcement and steel bars for the bottom layer of deck reinforcement. Characteristics of the four bridges selected for the field investigation are given in Table 2.2.1. Table 2.2.1: Characteristics of Bridges in Field Investigation (Frosch et al., 2006) Bridge No. of Spans Length of Spans (ft.) Girder Type Abutment Type Skew (deg.) I 65 over SR 25 2 76, 76 Steel Integral 25 SR 18 over I 65 2 123, 123 Steel Semi-Integral 30 SR 23 over US 20 2 101, 101 Precast/Prestressed Concrete Integral 11 Thayer Rd over I 65 5 39, 63, 77, 63, 39 Steel Rockers 5 The reinforcement selected in the bridge decks provided a range of longitudinal bar spacing values, between 6 and 18 in. The reinforcement size consisted of No. 4 or No. 5 bars. To facilitate the development of design recommendations related to crack control, the SR 18 bridge was constructed with different reinforcement details in the eastern and western spans, as illustrated in Table 2.2.2.

11 Table 2.2.2: Characteristics of deck reinforcement in field investigation (Frosch et al., 2006) Bridge Type of Reinforcement Bar Size Longitudinal Reinforcement Spacing (in.) Top Bottom Top Bottom Top Bottom I 65 over SR 25 Steel Steel 4 5 12 12 SR 18 over I 65 (Purdue Span) Steel Steel 4 4 6 6 SR 18 over I 65 (AASHTO Span) Steel Steel 5 5 18 12 SR 23 over US 20 Steel Steel 5 5 6 6 Thayer Rd over I 65 GFRP Steel 5 5 6 9 The Purdue span of the SR 18 bridge contained the reinforcement detail proposed by Frosch et al. for crack control, while the AASHTO Span was designed to conform to the requirements in the Standard Specifications for Highway Bridges, 16th Edition (AASHTO 1997) and the INDOT Standard Specifications dated 1999 (INDOT 1999). Visual crack mapping, beginning as early as 15 days after the placement of the CIP deck with subsequent mappings completed as late as 799 days after the placement of the CIP deck, was utilized in conjunction with measured strains to evaluate the performance of each bridge deck. Table 2.2.3 compares the crack statistics for the two spans of SR18, as well as the results of the Thayer Road bridge, which had five shorter spans. Both of these structures were wet-cured for seven days. As expected, for SR18 with the two different reinforcement details, the closer-spaced reinforcement tended to produce smaller cracks in greater numbers. The Purdue span was observed to have the most number of cracks; however the average crack width was the smallest, with a magnitude of 0.007 in., which was 30 percent less than the average crack width measured in the AASHTO span. Furthermore, less variation was observed in the crack widths, with a standard deviation of 0.005 in. compared to 0.008 in. for the AASHTO and Thayer Road spans. Finally, the maximum crack width observed in the Purdue span was 28 and 31 percent smaller than that measured in the AASHTO and Thayer Road spans, respectively. The approximately 30 percent improvement in both the average and maximum crack widths, in combination with the reduction in variability among the crack widths, suggested that the design parameters utilized to develop the Purdue span in the SR 18 bridge should be considered as a superior alternative, in terms of performance, to the AASHTO (1997) and INDOT (1999) design parameters. Furthermore, the authors suggested that the performance of the Thayer Road spans, which was similar to that observed in the AASHTO span, was due to increased slip between the concrete and FRP reinforcement.

12 Table 2.2.3: Comparison of crack width statistics (Frosch et al., 2006) Crack Statistics SR 18 Thayer Rd AASHTO Purdue Number of Cracks (Total) 15 22 19 Number of Cracks (per 100 ft.) 12.2 17.9 6.8 Average Crack Widths (in.) 0.010 0.007 0.010 Standard Deviation (in.) 0.008 0.005 0.008 Maximum Crack Width (in.) 0.025 0.018 0.026 Variance 3.11 x 10-5 2.53 x 10-5 5.56 x 10-6 In addition to the experimental results obtained during the study, the researchers completed a numerical parametric study to further investigate the expected effects of various bridge design characteristics on the performance of bridge decks. The finite element method (FEM) model was developed using ANSYS (SAS Inc. 2004) and was calibrated using the results of the Radabaugh (2001) study to investigate the effects of restrained shrinkage. Radabaugh (2001) constructed two laboratory specimens to study the effects of shrinkage restraint due to stay-in-place (SIP) deck pans. The specimens were 9 ft. by 9 ft. square, and represented the positive moment region of the bridge decks in the I-65 bridge span over SR 25. The specimens were developed in an effort to investigate the level of restraint based on the type of formwork, where one specimen had SIP steel deck forms, and the second was constructed with plywood forms and two layers of 10-mil Teflon as a barrier between the plywood and concrete. The second specimen was expected to provide little shrinkage restraint and therefore was developed to represent a state of free-shrinkage. Four layers of solid elements were selected to model the deck, which provided insight into the behavior of the deck through its depth, while reinforcement was discretely modeled using link elements. The materials were modeled as linear elastic, with the elastic modulus of the steel reinforcement and concrete of 29,000 ksi and 3,610 ksi, respectively. Uniform and linear concrete shrinkage profiles were considered, which were developed from the literature (Radabaugh, 2001 and Blackman, 2002). The calibrated FEM model was utilized to complete a numerical parametric study to investigate the effects of a range of variables on cracking in bridge decks, outlined in Table 2.2.4. The variables considered included reinforcement type, size, and spacing, the girder superstructure depth and spacing, and the thickness and material strength of the CIP deck.

13    Table 2.2.4: Range of variables considered in parametric study (Frosch et al., 2006)  Variable  Range  Control  Reinforcement  Area (Bar Size)  #3‐#7  #5  Top Mat Spacing (in.)  3‐18  12  Bottom Mat Spacing (in.)  3‐18  12  Reinforcement Type  Steel, FRP  Steel  Girder  Depth (in.)  12‐51  51  Spacing (ft.)  6‐10  7  Deck  Thickness (in.)  6‐12  8  f'c (psi)  3,000‐10,000  4,000    The results of the parametric study generally corroborated the results anticipated by the researchers,  such as increasing the area of reinforcement in a section tended to reduce the crack widths and  reinforcement stresses under a given load. The effects of reinforcement spacing were investigated by  varying the spacing while simultaneously maintaining a constant reinforcement ratio among models.  Using this method, the reinforcement spacing was observed to control two aspects of the behavior of  the models: (1) the crack width was observed to be larger as the reinforcement spacing increased, and  (2) the variation in the crack width between adjacent reinforcement was highly dependent on the bar  spacing (i.e., an “unzipping” effect was observed between adjacent reinforcing bars).   For example, in the case of uniform shrinkage, with a steel reinforcement ratio of 0.65 percent, the  following results were obtained. For the largest reinforcement spacing (i.e., 18 in.) considered in the  study, the maximum difference between the crack width near a bar and half the distance to an adjacent  bar was approximately 1.5 mils. Furthermore, the unzipping effect was dependent on the 3 in. mesh  size, which was evinced by the fact that a change in crack width of approximately 1/2 mil was observed  between measurements near and away from a bar in the 6 in. spacing model, while virtually no variation  in the crack width was observed between adjacent bars in the model representing a 9 in. reinforcement  spacing. A comparison of the performance of the different models was made by considering the  maximum crack widths in each model, which subsequently corresponded to the crack width measured  equidistant from adjacent reinforcement if a variation was observed between bars. A maximum crack  width of approximately 2.4 mils was observed in the model with a 3 in. reinforcement spacing, while a  maximum crack width of approximately 6.3 mils was observed in the model with an 18 in. reinforcement  spacing; where the increase in the crack widths was observed to be linearly proportional to the bar  spacing between these limits.  The authors suggested that if no slip is assumed to occur between the reinforcement and concrete, then  when cracking occurs, the tensile stresses previously restrained by the concrete will be transferred to  the reinforcement. If the concrete tensile strength is taken to be 6√f’c , then the stress in the  reinforcement upon cracking can be determined as shown in Eqn. (2.2.1).  ௦݂ ൌ ଺ඥ௙೎ఘ೒                                       (2.2.1) 

14    where fc is the concrete compressive strength at the initiation of cracking in psi (and the units of fc are  in psi), and ρg is the gross reinforcement ratio. The researchers plotted the average reinforcement  stresses versus the ratio √fc / ρg, and found that a linear fit yielded a line with a slope of approximately 3,  indicating that the average stress in the reinforcement bridging the crack was numerically determined to  be about half of what would be expected if no slip occurred. The slip length utilized in the model was 2  in.  Furthermore, the authors discovered that, on average, the reinforcement stresses increased by  approximately a factor of two between the simulated initial and final shrinkage states, where the initial  state was when cracking first occurred (6√fc) and the final was considered to be after a free shrinkage  “load” of 1000 µε was applied.  The researchers provided a recommended gross reinforcement ratio for crack control reinforcement  based on the results from the FEM parametric study, outlined above. Because the concrete compressive  strength at the time of cracking was unlikely to be known by the designer, the authors conservatively  substituted the 28‐day concrete compressive strength in the calculation. The recommended  reinforcement ratio proposed by Frosch et al. (2006) is given in equation 2.2.2).  ߩ௚ ൌ ଺ඥ௙೎ᇱ௙೤                                                   (2.2.2)  Furthermore, the reinforcement spacing required to restrain crack growth was developed during the  study and based on previous work. The reinforcement spacing, s, for Grade 60 bars is given as  ݏ ൑ 9 ቀ2.5 െ ௖೎ଶ ቁ ൑ 9 ݅݊.                                  (2.2.3)  where s is the reinforcement spacing (in.) and cc is the depth of concrete cover measured from the  extreme tensile fiber of the concrete to the face of reinforcement (in.).    2.3. Horizontal Shear Capacity of Composite Concrete Beams without Ties, Naito et al.,  2006  Composite precast construction increases the efficiency of precast sections because it may reduce the  section depth, as well as increase the load capacity, and the geometric stiffness of the system. A  composite concrete system must be designed such that the horizontal compression and tension forces  developed in the system can be transferred between the precast and CIP interface. Two design  documents for concrete construction in the United States, AASHTO (2010) and ACI (318‐08), include  design parameters for horizontal shear transfer. ACI (318‐08) allows for the design of composite sections  without the use of horizontal shear reinforcement, and permits a maximum factored horizontal shear  stress of 80 psi to be developed by a clean, roughened, surface with no ties crossing the shear plane.  AASHTO (2010) requires that horizontal shear reinforcement be present in all sections in which  composite action is to be achieved; however AASHTO provides a cohesion factor that may be  interpreted, for the purpose of discussion, as a maximum allowable horizontal shear strength (Naito et  al., 2006). The 2006 AASHTO LRFD design specification provided a cohesion factor of 100 psi for a clean,  intentionally roughened surface to an amplitude of 1/4 in., while AASHTO (2010) increased the cohesion  factor for a clean, intentionally roughened surface to 240 psi; which was not published at the time of  designing the NCHRP 10‐71 Concept 1 laboratory bridge specimen. Furthermore, research by Naito et al.  (2006) suggests that sections without vertical ties for horizontal shear reinforcement can develop  adequately large and reliable horizontal shear capacities.  

15    The calculation of the horizontal shear demand in a section can be completed using many different  methods. The researchers investigated three of these methods: (1) global force equilibrium, (2)  simplified elastic beam behavior, and (3) classic elastic methods (Naito et al., 2006). Global force  equilibrium considers a change in the compression force resultant over a distance along the length of  the beam, l. The horizontal shear strength is simply the change in the compression force divided by the  horizontal shear area, or bv *l, where bv is the width of the horizontal shear plane. For simply‐supported  systems, the horizontal shear stress can be calculated using this method by dividing the total  compressive force at midspan by half of the span length and the width of the horizontal shear plane. The  second method considers flexural beam theory, which equates the horizontal shear demand to the  vertical shear acting at a given section and is calculated as V/(bv d), where d is defined differently in ACI  and AASHTO. ACI (318‐08) defines d as the distance from the centroid of the longitudinal tension  reinforcement to the extreme compression fiber of the section, while AASHTO (2010) defines d as the  distance from the longitudinal tension reinforcement to the mid‐thickness of the slab. The third method  presented by the authors, the classic elastic method, considers the section properties of the beam;  however, the method is applicable to uncracked sections only. The horizontal shear stress calculated  using the classical elastic method is illustrated in Eqn. (2.3.1).  ݒ௛ ൌ ܸܳ ܫܾ௩ൗ                                                             (2.3.1)    Naito et al. (2006) investigated the horizontal shear stress developed in composite girders without  horizontal shear reinforcement. The researchers completed an experimental program with 19 test  specimens to investigate the effects of surface roughness, concrete strength, and loading type (i.e.,  point loading versus uniform loading). Four variations in the surface roughness were considered,  including as‐placed (A), broom finish (B), 1/4 in. rake (R), and sheepsfoot (Sh). The concrete compressive  strength in the flange was varied within a wide range of reasonably expected values, with measured  concrete strengths of 3.11, 5.67, 8.75, and 9.71 ksi. Two‐point and 5‐point loading configurations were  utilized during the study to simulate point loading and uniform loading, respectively. Load was applied  through 12 in. neoprene bearing pads in an effort to reduce the localized bearing stress near the applied  load. A summary of the research parameters and specimen characteristics is outlined in Table 2.3.1  (Naito et al., 2006).  The researchers provided instrumentation to determine the horizontal shear stress at the interface as  well as to measure the slip that may occur along predicted failure planes. Two horizontally oriented  strain gages were mounted on the surface of the CIP flange along the depth, to allow for the  measurement of the horizontal shear stress with the use of known stress‐strain relationships for the CIP  concrete.     

16 Table 2.3.1: Research parameters and specimen characteristics considered during the study (Naito et al., 2006) Beam Interface Finish Loading Method Interface Width (in.) Web Steel Area (in.2) Flange Strength (ksi) Effective Prestress (ksi) 1 As-Placed 5 Point 5 0.2 5.67 141.3 2 Broom 5 Point 5 0.2 5.67 142.1 3 Monolithic 5 Point 5 0.0 9.71 139.9 4 Rake 5 Point 5 0.2 3.11 141.5 5 Rake 5 Point 5 0.2 5.67 143.1 6 Rake 5 Point 5 0.2 8.75 140.3 7 Sheepsfoot 5 Point 5 0.2 5.67 140.3 8 As-Placed 2 Point 2 0.2 5.67 140.2 9 As-Placed 2 Point 2 0.2 5.67 140.1 10 Broom 2 Point 2 0.2 5.67 140.2 11 Monolithic 2 Point 2 0.0 9.71 140.2 12 Monolithic 2 Point 2 0.0 9.71 140.2 13 Rake 2 Point 2 0.2 3.11 140.2 14 Rake 2 Point 2 0.2 3.11 140.2 15 Rake 2 Point 2 0.2 5.67 140.2 16 Rake 2 Point 2 0.2 5.67 140.2 17 Rake 2 Point 2 0.0 8.75 140.2 18 Rake 2 Point 2 0.0 8.75 140.2 19 Smooth 2 Point 2 0.2 5.67 140.2 Relatively large horizontal shear stresses were measured during both the 2-point and 5-point loading cases, and were always above the 80 psi allowed for sections without horizontal shear ties as specified by ACI (318-08) and the 240 psi cohesion value provided by AASHTO (2010). During the 5-point load tests, the horizontal shear stress was measured at the first indication of cracking in the section, generally flexure shear cracking, in an effort to investigate the shear stress developed before cracking, which would generally be the case for a service condition, where cracking is generally avoided. The results of the 5-point load tests are illustrated in Table 2.3.2, with an average horizontal shear stress at cracking of 340.2 psi using the classic elastic method of calculation. The smallest average horizontal shear stress was 275.4 psi, calculated using the simplified elastic method provided by ACI (318-08).

17 Table 2.3.2: Horizontal shear stress at cracking (psi) during 5-point load tests (Naito et al., 2006) Beam Interface Finish Elastic VQ/Ibv ACI V/bvdp AASHTO V/bvdv 1 As-Placed 341.1 276.0 334.5 2 Broom 341.1 276.0 334.5 3 Monolithic 350.1 280.0 339.4 4 Rake 321.6 266.0 322.4 5 Rake 341.1 276.0 334.5 6 Rake 345.6 278.0 337.0 7 Sheepsfoot 341.1 276.0 334.5 Average 340.2 275.4 333.9 The horizontal shear stresses developed during the 2-point load tests were calculated at the ultimate capacity of each specimen, and were subsequently considerably larger than those measured during the five-point load configurations. As illustrated in Table 2.3.3, the average horizontal shear stress measured at ultimate ranged from 828.4 psi to 1022.3 psi, depending on the method of calculation.

18 Table 2.3.3: Horizontal shear stress at ultimate (psi) during two-point load tests (Naito et al., 2006) Beam Interface Finish From Strain C/Lbv Elastic VQ/Ibv ACI V/bvdp AASHTO V/bvdv 8 As-Placed 482.2 863.2 698.4 846.6 9 As-Placed 814.0 1060.8 860.7 1043.2 10 Broom 915.6 993.2 804.6 975.3 11 Monolithic 1075.0 1067.0 848.4 1028.4 12 Monolithic 1288.0 1248.1 981.1 1189.2 13 Rake 639.0 850.6 703.7 852.9 14 Rake 1182.0 1015.1 840.2 1018.4 15 Rake 1348.0 1001.0 811.1 983.2 16 Rake 1245.0 1165.9 948.1 1149.2 17 Rake 1054.0 1141.3 934.3 1132.5 18 Rake 1194.0 1073.6 873.1 1058.3 19 Smooth 787.4 787.9 637.5 772.7 Average 1002.0 1022.3 828.4 1004.1 The experimental results indicated that composite girders without horizontal shear reinforcement can develop significant horizontal shear stresses. The authors suggested that, with an average horizontal shear stress of 340.2 psi at the service state, which was more than three times the least conservative design estimate, the requirement for horizontal shear reinforcement in all sections by AASHTO (2005) be waived. 2.4. AASHTO (2007) Bursting Design Requirements The AASHTO (2007) specifications required vertical reinforcement to resist four percent of the prestressing force in the end regions of precast beams over a horizontal distance of h/4 from the face, where h is the section height. The provisions are for pretensioned anchorage regions and do not provide any distinction based on cross section. The AASHTO (2007) specifications resulted in the requirement of a large amount of vertical steel over a short distance for wide shallow members, such as the shallow inverted T sections used in PCSSS bridges. This created congestion of the vertical reinforcement in the end zones and provided an impetus for a review of the requirements and the end zone force demands for the case of shallow inverted-T sections. In reviewing the literature, the origin of the AASHTO bursting requirements likely resulted from a Marshall and Mattock (1962) study on horizontal end zone cracking of pretensioned I-girders. The 1961 Interim AASHTO Specifications introduced bursting reinforcement requirements (in the AASHTO (2010) LRFD Bridge Design Specifications this is labeled “splitting resistance”) for pretensioned I-beams stating, “Vertical stirrups acting at 20,000 psi to resist 4 percent of the prestressing force shall be placed within

19 d/4 from the end of the beam with the end stirrup placed as close to end of beam as practicable,” where d is the effective depth. By 1969, the above requirements were not limited to just I-beams, and instead were made applicable to end zones of all pretensioned beams. The AASHTO LRFD Bridge Design Specifications 1st edition (1994) had the same requirements as the 1969 version, but the 2nd edition (1998) stated that the vertical steel should be placed within h/5 from the end face, where h is the height of the member. This specification was changed to h/4 from the end face in the 3rd edition (2004). The only difference between the 1969 provisions and the 4th edition (2007) of the AASHTO LRFD Bridge Design Specifications was that vertical steel could be placed within h/4 from the end face, rather than d/4. The 2008 Interim specifications relaxed the requirements for wide-shallow sections, by allowing the designer to spread the end zone reinforcement, termed “splitting” reinforcement over a larger distance. In the case of pretensioned solid or voided slabs, the designer can substitute the section width for “h,” rather than using the section depth for “h.” Because the original AASHTO provisions were likely developed as a result of experimental testing performed on I-beams by Marshall and Mattock (1962), it is not clear whether the AASHTO requirements, developed for I-beam members, are applicable to other cross sections, particularly slab- span systems. 2.4.1. Bursting, Splitting and Spalling Stresses The end regions of prestressed concrete sections have complicated states of stress. Beam theory is not applicable in the end regions of prestressed concrete beams because the longitudinal strain is not linearly distributed through the depth of the cross section due to the introduction of the prestress force. As the prestress force distributes through the section, various zones of compression and tension are created over a region assumed to extend a distance approximately equal to the depth of the section, in accordance with St. Venant’s principle. The tensile stresses that develop in this region have been identified as bursting or spalling stresses, depending on their location. Splitting stresses are circumferential tensile stresses resulting from the radial compressive stresses caused by bond. Splitting stresses occur along the transfer length and locally around the prestressing strand. Bursting and spalling stresses are vertical tensile stresses that develop in the concrete from the distribution of the prestressing force as the force is transferred from the prestressing strand to the concrete at the ends of the member and can result in cracking at different locations in the beam. Spalling stresses are a maximum at the end face of the member, typically near the centroid of the section and result in cracking at the end face, which can propagate further into the member (Gergely et al., 1963). Bursting stresses occur along the line of the prestressing force, beginning a few inches into the beam and extending through the transfer length. Bursting and splitting stresses can result in strand slippage as cracking at the strand can eliminate bond between the strand and the concrete. Figure 2.4.1 shows the location of the bursting and spalling stresses in the end region of a section. Note that the AASHTO (2007) provisions (incorrectly labeled as bursting provisions) and the AASHTO (2010) provisions (incorrectly labeled as splitting provisions) are intended to resist spalling forces (i.e., vertical tensile forces occurring near the end face at the centroid of the member). AASHTO (2010) does not explicitly include provisions for vertical steel to resist bursting and splitting forces (i.e., vertical forces occurring along the line of the prestress in the transfer length region); however the AASHTO confinement reinforcement provisions fill this role. Distribution of tensile stresses in the end region depends on the eccentricity of the prestressing force in the member, where eccentricity is the distance between the centroid of the prestressing force and the centroid of the member. In a centrally loaded member, forces distribute symmetrically through the

20 vertical member height. Small spalling forces develop on the end face and a bursting force starts further into the member, as shown in Figure 2.4.1-a. In members with large eccentricity, there is greater area above the prestressing force in which the stresses need to distribute. This allows the prestressing force to spread through a greater vertical distance, making the curvature of the flow of stresses greater, creating a larger spalling force near the end region, as shown in Figure 2.4.1-b. This is consistent with previous experimental work which related the maximum tensile stress location to eccentricity. Gergely’s (1963) study on post-tensioned I-girders and rectangular sections found members with small eccentricities had maximum tensile stresses in the bursting zone and members with large eccentricities had maximum tensile stresses in the spalling zone. Hawkins (1960) corroborated Gergely’s findings and also found as eccentricity increased so did the magnitude of maximum tensile stress in the spalling zone. The 12 in. deep precast inverted tee sections of the laboratory bridge specimen had slight eccentricities of 2.4 in. for the 3 in. flanged section and 2.5 in. for the 5.25 in. flanged section. Because it was unclear whether this shallow section would have a maximum tensile force in the bursting or spalling zone, tensile stresses in both zones were investigated. (a) Centrally prestressed member (b) Eccentrically prestressed member Figure 2.4.1: Effects of eccentricity of distribution of compression and spalling forces 2.4.2. Stresses in End Regions of Post-Tensioned and Pretensioned Sections Many previous studies investigated the effects of bursting and spalling stresses in post-tensioned members (Gergely et al., 1963; Hawkins, 1960). Only a few examined pretensioned members. The primary difference between post-tensioned and pretensioned systems is the means of introducing the distribution of the prestress force into the members. Post-tensioned systems transfer stress from the steel to the concrete at the end face, through bearing at the end plate, which spreads the force over the end of the member. In pretensioned systems, where the transfer of stress occurs through bond, the stress transfer is often assumed to vary linearly or uniformly until full transfer occurs. The distance needed to transfer the stress from steel to concrete is defined as the transfer length. Research has shown that tensile stresses in end zones are affected by the transfer length (Base, 1958). Longer transfer lengths in pretensioned systems result in smaller bursting and spalling stresses. Shorter transfer lengths concentrate the transfer of forces, which result in larger bursting and spalling stresses, more similar to the case of post-tensioned systems (Uijl, 1983). Many theories developed from post-

21 tensioned experiments can provide conservative estimates of the spalling and bursting stresses in pretensioned members, because they simulate the case of a very short transfer length.